AWS D1.1/D1.1M pdf download

General requirements2. Design of welded connectios3. Prequalification of WPSs4. Stud welding8. General requirementsC Design of welded connectionsC Prequalification of WPSsC The lateral force-resisting system in buildings may consist of a vertical truss.

This is referred to as a braced frame and the connections of the diagonal braces to the beams and columns are the bracing connections. Figure 2. For the bracing system to be a true truss, the bracing connections should be concentric, that is, the gravity axes of all members at any joint should intersect at a single point. If the gravity axes are not concentric, the resulting couples must be considered in the design of the members.

Consider the bracing connection of Fig. The brace load is kips, the beam shear is 10 kips, and the beam axial force is kips. The design of this connection involves the design of four separate connections. These are 1 the brace-to-gusset connection, 2 the gusset-to-column connection, 3 the gusset-to-beam connection, and 4 the beam-to-column connection.

A fifth connection is the connection on the other side of the column, which will not be considered here. Brace-to-gusset: This part of the connection is designed first because it provides a minimum size for the gusset plate which is then used to design the gusset-to-column and gusset-to-beam connections. Providing an adequate load path involves the following limit states: a.

Therefore, try 12 bolts each side of the connection. The resistances of the individual bolts are then summed to determine a capacity for the bolt group. The bolt shear strength has already been established as However, since the connection is to be designed as slip critical, the slip resistance will govern. Because of the fillet weld of the gusset to the beam flange, this part of the Whitmore section is not ineffective, that is, load can be passed through the weld to be carried on this part of the Whitmore section.

The effective length of the Whitmore section is thus The gusset buckling length is, from Fig. Thus 0. Per angle, This completes the design checks for the brace-to-gusset connection. All elements of the load path, which consists of the bolts, the brace web, the gusset, and the connection angles, have been checked. The remaining connection interfaces require a method to determine the forces on them. The force distributions for this method are shown in Fig. From the design of the brace-to-gusset connection, a certain minimum size of gusset is required.

This is the gusset shown in Fig. Usually, this gusset size, which is a preliminary size, is sufficient for the final design. From Fig. From the geometry given in Fig. Gusset-to-column: The loads are kips shear and 0 kip axial. Fillet weld of clip angles to gusset: The length of this clip angle weld is 28 in. See Table 1. Gusset-to-beam: The loads are kips shear and kips axial. The length of the gusset is The 1-in snip can be ignored with negligible effect on the stress.

Gusset stresses: b. The factor 1. Even though the stress in this weld is calculated as being uniform, it is well known that there will be local peak stresses, especially in the area where the brace-to-gusset connection comes close to the gusset-to-beam weld. An indication of high stress in this area is also indicated by the Whitmore section cutting into the beam web. Also, as discussed later, frame action will give rise to distortional forces that modify the force distribution given by the UFM.

Thus which is reasonable. Note that this length can be longer than the gusset-to-beam weld, but probably should not exceed about half the beam span. The vertical component can cause beam web yielding and crippling. The physical situation is closer to that at some distance from the beam end rather than that at the beam end.

Beam to column: The loads are kips axial, the specified transfer force and a shear which is the sum of the nominal minimum beam shear of 10 kips and the vertical force from the gusset-to-beam connection of kips. In this connection, since the bolts also see a tensile load, there is an interaction between tension and shear that must be satisfied.

If V is the factored shear per bolt, the design tensile strength is This formula is obtained by inverting Specification formula Ja. Let the end plate be 11 in wide. The edge distance at the top and bottom of the end plate is 1. In the above expression, p is the tributary length of end plate per bolt. For the bolts adjacent to the beam web, this is obviously 4 in. Weld of beam to end plate: All of the shear of kips exists in the beam web before it is transferred to the end plate by the weld of the beam to the end plate.

It is a tenet of all gusset plate designs that it must be able to be shown that the stresses on any cut section of the gusset do not exceed the yield stresses on this section. Now, once the resultant forces on the gusset horizontal and vertical sections are calculated by the UFM, the resultant forces on any other cut section, such as section a-a of Fig.

The traditional approach to the determination of stresses, as mentioned in many books Blodgett, ; Gaylord and Gaylord, ; Kulak et al. It is well known that these are not correct for gusset plates Timoshenko, They are recommended only because there is seemingly no alternative. Actually, the UFM, coupled with the Whitmore section and the block shear fracture limit state, is an alternative as will be shown subsequently.

Applying the slender member formulas to the section and forces of Fig. Since 9. Since At this point, it appears that the design is unsatisfactory i. But consider that the normal stress exceeds yield over only about 11 in of the in-long section starting from point A. It cannot be achieved in an elastic—perfectly plastic material with a design yield point of What will happen is that when the design yield point of At this time, the plate will fail by unrestrained yielding if the applied loads are such that higher stresses are required for equilibrium.

To conclude on the basis of What must be done is to see if a redistributed stress state on the section can be achieved which nowhere exceeds the design yield stress. Note that if this can be achieved, all AISC requirements will have been satisfied.

The AISC specifies that the design yield stress shall not be exceeded, but does not specify the formulas used to determine this. The shear stress fv and the axial stress fa are already assumed uniform. Only the bending stress fb is nonuniform. To achieve simultaneous yield over the entire section, the bending stress must be adjusted so that when combined with the axial stress, a uniform normal stress is achieved. To this end, consider Fig. Here the bending stress is assumed uniform but of different magnitudes over the upper and lower parts of the section.

Note that this can be done because M of Fig. This being the case, there is no reason to assume that the bending stress distribution is symmetrical about the center of the section. Considering the distribution shown in Fig. Equating the couple M of Fig.

Since at all points of the section, the design yield stress is nowhere exceeded and the connection is satisfactory. It was stated previously that there is an alternative to the use of the inappropriate slender beam formulas for the analysis and design of gusset plates.

The UFM performs exactly the same analysis on the gusset horizontal and vertical edges, and on the associated beam-to-column connection. It is capable of producing forces on all interfaces that give rise to uniform stresses. Each interface is designed to just fail under these uniform stresses.

Therefore, true limit states are achieved at every interface. For this reason, the UFM achieves a good approximation to the greatest lower bound solution closest to the true collapse solution in accordance with the lower bound theorem of limit analysis. The UFM is a complete departure from the so-called traditional approach to gusset analysis using slender beam theory formulas.

It has been validated against all known full-scale gusseted bracing connection tests Thornton, , b. It does not require the checking of gusset sections such as that studied in this section section a-a of Fig.

The analysis at this section was done to prove a point. But the UFM does include a check in the brace-to-gusset part of the calculation that is closely related to the special section a-a of Fig. This is the block shear rupture of Fig. The block shear capacity was previously calculated as kips.

Comparing the block shear limit state to the special section a-a limit state, a reserve capacity in block shear is found, and the reserve capacity of the special section , which shows that block shear gives a conservative prediction of the capacity of the closely related special section.

A second check on the gusset performed as part of the UFM is the Whitmore section check. With these two limit states, block shear rupture and Whitmore, the special section limit state is closely bounded and rendered unnecessary. The routine calculations associated with block shear and Whitmore are sufficient in practice to eliminate the consideration of any sections other than the gusset-to-column and gusset-to-beam sections. This connection is shown in Fig.

The completed design is shown in Fig. In this case, because of the high specified beam shear of kips, it is proposed to use a special case of the UFM which sets the vertical component of the load between the gusset and the beam, VB, to zero.

Brace-to-gusset connection: a. Weld: The brace is field welded to the gusset with fillet welds. From the geometry of the gusset and brace, about 17 in of fillet weld can be accommodated.

The average thickness of 0. The thickness at the toe of the fillet is 0. The Whitmore section extends into the column by 5. The column web is stronger than the gusset since 1. The Whitmore also extends into the beam web by 6. Since the brace force can be tension or compression, compression will control. This completes the brace-to-gusset part of the design. Before proceeding, the distribution of forces to the gusset edges must be determined.

The same results can be obtained formally with the UFM by setting and proceeding as follows. Continuing This couple is clockwise on the gusset edge. It can be seen that these gusset interface forces are the same as those obtained from the simpler method. Gusset-to-column connection: The loads are kips shear and kips axial.

The reason for this lies in the physical behavior of slip-critical connections. Connection shear, Vu, is carried by the faying surface through friction—rather than by the bolt shank—until slip occurs.

Once slip occurs, bearing interaction Equation Ja from the Specification and the prying action model as shown in the Manual must be used Thornton, Following the notation of the Manual.

In this case bending in the plate governs. In addition to the prying check, the end plate should be checked for gross shear, net shear, and block shear. These will not govern in this case. Checks on column web: 1 Web yielding under normal load Hc : 2 Web crippling under normal load Hc : 3 Web shear: The horizontal force, Hc, is transferred to the column by the gusset-to-column connection and back into the beam by the beam-to-column connection.

Gusset-to-beam connection: The loads are kips shear and a kips-in couple. Weld of gusset-to-beam flange: Since The 1. This method does not give an indication of peak and average stresses, but it will be safe to use the ductility factor. The shear is thus This shear is applied to the flange as a transverse load over 15 in of flange.

It does not reach the beam-to-column connection where the beam shear is kips. However, the AISC book on connections AISC, addresses this situation and states that because of frame action distortion , which will always tend to reduce Hc, it is reasonable to use the larger of Hc and A as the axial force.

Thus the axial load would be kips in this case. It should be noted however that when the brace load is not due to primarily lateral loads frame action might not occur.

As mentioned earlier the slip-critical strength criterion in used. Since all of the bolts are subjected to tension simultaneously, there is interaction between tension and shear. The reduced tensile capacity is Prying action is now checked using the method and notation of the AISC Manual of Steel Construction , pages through Check 1. Since 3.

These will not control in this case. The following method can be used when tf and tp are of similar thicknesses. Because of the axial force, the column flange can bend just as the clip angles. A yield-line analysis derived from Mann and Morris can be used to determine an effective tributary length of column flange per bolt. The yield lines are shown in Fig. Thus, Using peff in place of p, and following the AISC procedure, Note that standard holes are used in the column flange.

The method of bracing connection design presented here, the uniform force method UFM , is an equilibrium-based method. Every proper method of design for bracing connections, and in fact for every type of connection, must satisfy equilibrium. The set of forces derived from the UFM, as shown in Fig. If it is assumed that the structure and connection behave elastically an assumption as to constitutive equations and that the beam and the column remain perpendicular to each other an assumption as to deformation—displacement equations , then an estimate of the moment in the beam due to distortion of the frame frame action Thornton, is given by With and This moment MD is only an estimate of the actual moment that will exist between the beam and column.

The actual moment will depend on the strength of the beam-to-column connection. The strength of the beam-to-column connection can be assessed by considering the forces induced in the connection by the moment MD as shown in Fig. The distortional force FD is assumed to act as shown through the gusset edge connection centroids. If the brace force P is a tension, the angle between the beam and column tends to decrease, compressing the gusset between them, so FD is a compression. If the brace force P is a compression, the angle between the beam and column tends to increase and FD is a tension.

For the elastic case with no angular distortion It should be remembered that these are just estimates of the distortional forces. The actual distortional forces will be dependent also upon the strength of the connection. Compare, for instance, HD to Hc. Hc is kips tension when HD is kips compression. The strength of the connection can be determined by considering the strength of each interface, including the effects of the distortional forces.

The following interface forces can be determined from Figs. Thus, the design shown in Fig. Note also that NBC could have been set as The NBC value of kips is used to cover the case when there is no excess capacity in the beam-to-column connection. Now, the gusset-to-beam and gusset-to-column interfaces will be checked for the redistributed loads of Fig. Gusset to Beam. Gusset stresses: 2. No ductility factor is used here because the loads include a redistribution. Gusset to Column. This connection is ok without calculations because the loads of Fig.

It has been shown that the connection is strong enough to carry the distortional forces of Fig. In general, the entire connection could be designed for the combined UFM forces and distortional forces, as shown in Fig. This set of forces is also admissible. The UFM forces are admissible because they are in equilibrium with the applied forces.

The distortional forces are in equilibrium with zero external forces. Under each set of forces, the parts of the connection are also in equilibrium. Therefore, the sum of the two loadings is admissible because each individual loading is admissible. A safe design is thus guaranteed by the lower bound theorem of limit analysis.

The difficulty is in determining the distortional forces. The elastic distortional forces could be used, but they are only an estimate of the true distortional forces. The distortional forces depend as much on the properties of the connection, which are inherently inelastic and affect the maintenance of the angle between the members, as on the properties and lengths of the members of the frame.

The UFM produces a load path that is consistent with the gusset plate boundaries. For instance, if the gusset-to-column connection is to a column web, no horizontal force is directed perpendicular to the column web because unless it is stiffened, the web will not be able to sustain this force. This is clearly shown in the physical test results of Gross where it was reported that bracing connections to column webs were unable to mobilize the column weak axis stiffness because of web flexibility.

A mistake that is often made in connection design is to assume a load path for a part of the connection, and then to fail to follow through to make the assumed load path capable of carrying the loads satisfying the limit states.

Note that load paths include not just connection elements, but also the members to which they are attached. As an example, consider the connection of Fig. This is a configuration similar to that of Fig. This will be called the L weld method, and is similar to model 4, the parallel force method, which is discussed by Thornton In the example of Fig.

This results in free-body diagrams for the gusset, beam, and column as shown in Fig. Imagine how difficult it would be to obtain the forces on the free-body diagram of the gusset and other members if the weld were not uniformly loaded! Every inch of the weld would have a force of different magnitude and direction.

Note that while the gusset is in equilibrium under the parallel forces alone, the beam and the column require the moments as shown to provide equilibrium.

For comparison, the free-body diagrams for the UFM are given in Fig. These forces are always easy to obtain and no moments are required in the beam or column to satisfy equilibrium.

While the L weld method weld is very small, as expected with this method, now consider the load paths through the rest of the connection. The column web sees a transverse force of 80 kips. It should be noted that the yield-line pattern of Fig. That analysis assumed double curvature with a prying force at the toes of the end plate a distance a from the bolt lines.

But the column web will bend away as shown in Fig. Thus, single curvature bending in the end plate must be assumed, and the required end plate thickness is given by AISC The weld is already designed.

The beam must be checked for web yield and crippling, and web shear. The kip vertical load between the gusset and the beam flange is transmitted to the beam-to-column connection by the beam web. The doubler must start at a distance x from the toe, where 4. Therefore, a doubler of length 34 — The doubler thickness td required is 1. If some yielding before ultimate load is reached is acceptable, grade 36 plate can be used.

Beam to Column. The fourth connection interface the first interface is the brace-to-gusset connection, not considered here , the beam-to-column, is the most heavily loaded of them all. The 80 kips horizontal between the gusset and column must be brought back into the beam through this connection to make up the beam strut load of kips axial. This connection also sees the kips vertical load from the gusset-to-beam connection.

The reduced tension design strength is so use kips. If stiffeners are used, the most highly loaded one will carry the equivalent tension load of three bolts or The shear in the stiffener is Weld of Stiffener to Column Web.

Assume about 3 in of weld at each gage line is effective, that is 1. The load kips. The length of the weld is Additional Discussion. The kip horizontal force between the gusset and the column must be transferred to the beam-to-column connections. Therefore, the column section must be capable of making this transfer. These couples could act on the gusset-to-column and gusset-to-beam interfaces, since they are free vectors, but this would totally change these connections.

This will greatly reduce its capacity to carry kips in compression and is probably not acceptable. This completes the design of the connection by the L weld method. The final connection as shown in Fig. The column stiffeners are expensive, and also compromise any connections to the opposite side of the column web. The web-doubler plate is an expensive detail and involves welding in the beam k-line area, which may be prone to cracking AISC, Finally, although the connection is satisfactory, its internal admissible force distribution that satisfies equilibrium requires generally unacceptable couples in the members framed by the connection.

As a comparison, consider the design that is achieved by the UFM. The statically admissible force distribution for this connection is given in Fig. Note that all elements gusset, beam, and column are in equilibrium with no couples. Note also how easily these internal forces are computed. The final design for this method, which can be verified by the reader, is shown in Fig.

There is no question that this connection is less expensive than its L weld counterpart in Fig. As a final comment, a load path assumed for part of a connection affects every other part of the connection, including the members that frame to the connection. All of the bracing connection examples presented here have involved connections to the column using end plates or double clips, or are direct welded.

The UFM is not limited to these attachment methods. Figures 2. These connections are much easier to erect than the double-angle or shear plate type because the beams can be brought into place laterally and easily pinned. For the column web connection of Fig.

The connections shown were used on an actual job and were designed for the tensile strength of the brace to resist seismic loads in a ductile manner. The UFM can be easily generalized to this case as shown in Fig. It should be noted that this non-concentric force distribution is consistent with the findings of Richard , who found very little effect on the force distribution in the connection when the work point is moved from concentric to non-concentric locations.

In the case of Fig. In the case of a connection to a column web, this will be the actual distribution Gross, , unless the connection to the column mobilizes the flanges, as for instance is done in Fig. An alternate analysis, where the joint is considered rigid, that is, a connection to a column flange, the moment M is distributed to the beam and column in accordance with their stiffnesses the brace is usually assumed to remain an axial force member and so is not included in the moment distribution , can be performed.

Example Consider the connection of Sec. Using the data of Fig. This figure also shows the original UFM forces of Fig. The design of this connection will proceed in the same manner as shown in Sec. The UFM as originally formulated can be applied to trusses as well as to bracing connections.

After all, a vertical bracing system is just a truss as seen in Fig. But bracing systems generally involve orthogonal members, whereas trusses, especially roof trusses, often have a sloping top chord. Sindel D. Stickel BM. Toth K. Martin D. Phillips K. Miller, Chair ME. Gase, Vice Chair S. E, Anderson L. Clarke H. Gilmer M. Grieco C. Carbonneau JH. Kinney V. Kurwilla G. Martin S. Mattfield D. Medlock E. Mellinger RL.

Mertz R. E, Monson E. Alexander B. Anderson J. Cagle GL. Fox G. Hill L. Holdren W. Kinney, Chair J. Kinsey, Vice Chair M. E, Guse, 2 Vice Chair S. E, Anderson U. W, Aschemeier R. Clarke JA. Cochran IM. Davis PA. Furr H. Gilmer C.

Hayes: PT. Hilton N. Lindell G. Manin E. Mattficld JE, Mellinger I. Pearson, Jr E, Pennington R. Stachel K. Dunn J. J, Edwards G. Hill R. Holbert JH. Kiefer CA, Mankenberg. Houston, Chair U. Campbell A. D'Amico B. C, Hobson LW. Houston I. E, Koski D. Luciani CW. Makar S. Moran P. Wirtz P. B, Champney 3. Guili R. Schraft M. Grieeo, Vice Chair E. Bickford N. Choy R. Clarke D. Ferrell R. Fletcher P. Huckabee L. Kloiber C. Long P. Marslender J. Olson ILA. Both permit sealing the HSS ends with cap plates.

The above connections may be structurally efficient but they would generally be perceived as being unsuitable for architecturally exposed structural steel AESS. If bolted connections need to be hidden from view, cover plates — shaped to match the HSS — can obscure the bolted shear connection Figure 4.

Site bolting can still be performed and thin cover plates lightly screwed into place after. Bolted end-plate connections Figure 5 are another widely used method for bolting beyond the HSS member, whether the member is in tension or compression or in combination with bending.

Design procedures and design examples for axially loaded tension connections, for both rectangular bolted on two sides or on all four sides and round HSS, are given in AISC Design Guide No. These design methods entail a consideration of prying action on the bolts, which should be fully pre-tensioned. One advantage of this connection type is that it allows HSS of different size to be spliced together.

The connection can then be conservatively designed on the basis of the maximum effective tension load.